土木工程外文翻譯---鋼筋混凝土填充框架結構對拆除兩個相鄰的柱的響應

1、<p><b>　　本科畢業(yè)設計</b></p><p><b>　　外文文獻及譯文</b></p><p>　　文獻、資料題目：A comparative study of various </p><p>　　commercially available programs </p><p&

2、gt;　　in slope stability analysis</p><p>　　文獻、資料來源：Computers and Geotechnics</p><p>　　文獻、資料發(fā)表（出版）日期：2008.8.9</p><p><b>　　院 （部）：</b></p><p><b>　　專 業(yè)：

3、</b></p><p><b>　　班 級：</b></p><p><b>　　姓 名：</b></p><p><b>　　學 號：</b></p><p><b>　　指導教師：</b></p><

4、p><b>　　翻譯日期：</b></p><p>　　附件一，外文翻譯原文及譯文</p><p><b>　　1，文獻原文：</b></p><p>　　Response of a reinforced concrete infilled-frame structure to removal of two adja

5、cent columns</p><p>　　Mehrdad Sasani_</p><p>　　Northeastern University, 400 Snell Engineering Center, Boston, MA 02115, United States</p><p>　　Received 27 June 2007; received in rev

6、ised form 26 December 2007; accepted 24 January 2008</p><p>　　Available online 19 March 2008</p><p><b>　　Abstract</b></p><p>　　The response of Hotel San Diego, a six-sto

7、ry reinforced concrete infilled-frame structure, is evaluated following the simultaneous removal of two adjacent exterior columns. Analytical models of the structure using the Finite Element Method as well as the Applied

8、 Element Method are used to calculate global and local deformations. The analytical results show good agreement with experimental data. The structure resisted progressive collapse with a measured maximum vertical displac

9、ement of only one </p><p>　　c 2008 Elsevier Ltd. All rights reserved.</p><p>　　Keywords: Progressive collapse; Load redistribution; Load resistance; Dynamic response; Nonlinear analysis; Brittle

10、 failure</p><p>　　1. Introduction</p><p>　　As part of mitigation programs to reduce the likelihood of mass casualties following local damage in structures, the General Services Administration [1

11、] and the Department of Defense [2] developed regulations to evaluate progressive collapse resistance of structures. ASCE/SEI 7 [3] defines progressive collapse as the spread of an initial local failure from element to e

12、lement eventually resulting in collapse of an entire structure or a disproportionately large part of it. Following the approaches</p><p>　　2. Building characteristics</p><p>　　Hotel San Diego wa

13、s constructed in 1914 with a south annex added in 1924. The annex included two separate buildings. Fig. 1 shows a south view of the hotel. Note that in the picture, the first and third stories of the hotel are covered wi

14、th black fabric. The six story hotel had a non-ductile reinforced concrete (RC) frame structure with hollow clay tile exterior infill walls. The infills in the annex consisted of two wythes (layers) of clay tiles with a

15、total thickness of about 8 in (203 mm). Th</p><p>　　beams were present.</p><p>　　3. Sensors</p><p>　　Concrete and steel strain gages were used to measure changes in strains of beams

16、 and columns. Linear potentiometers were used to measure global and local deformations. The concrete strain gages were 3.5 in (90 mm) long having a maximum strain limit of ±0.02. The steel strain gages could measure

17、 up to a strain of ±0.20. The strain gages could operate up to a several hundred kHz sampling rate. The sampling rate used in the experiment was 1000 Hz. Potentiometers were used to capture rotation (integ</p>

18、<p>　　4. Finite element model</p><p>　　Using the finite element method (FEM), a model of the building was developed in the SAP2000 [8] computer program. The beams and columns are modeled with Bernoull

19、i beam elements. Beams have T or L sections with effective flange width on each side of the web equal to four times the slab thickness [5]. Plastic hinges are assigned to all possible locations where steel bar yielding c

20、an occur, including the ends of elements as well as the reinforcing bar cut-off and bend locations. The characteristics</p><p>　　The beams in the building did not have top reinforcing bars except at the end

21、regions (see Fig. 4). For instance, no top reinforcement was provided beyond the bend in beam A1–A2, 12 inches away from the face of column A1 (see Figs. 4 and 5). To model the potential loss of flexural strength in thos

22、e sections, localized crack hinges were assigned at the critical locations where no top rebar was present. Flexural strengths of the hinges were set equal to Mcr. Such sections were assumed to lose thei</p><p&

23、gt;　　The floor system consisted of joists in the longitudinal direction (North–South). Fig. 6 shows the cross section of a typical floor. In order to account for potential nonlinear response of slabs and joists, floors a

24、re molded by beam elements. Joists are modeled with T-sections, having effective flange width on each side of the web equal to four times the slab thickness [5]. Given the large joist spacing between axes 2 and 3, two re

25、ctangular beam elements with 20-inch wide sections are used betwe</p><p>　　equal to one-half of that of the gross sections [9].</p><p>　　The building had infill walls on 2nd, 4th, 5th and 6th fl

26、oors on the spandrel beams with some openings (i.e. windows and doors). As mentioned before and as part of the demolition procedure, the infill walls in the 1st and 3rd floors were removed before the test. The infill wal

27、ls were made of hollow clay tiles, which were in good condition. The net area of the clay tiles was about 1/2 of the gross area. The in-plane action of the infill walls contributes to the building stiffness and strength

28、and</p><p>　　Using the SAP2000 computer program [8], two types of modeling for the infills are considered in this study: one uses two dimensional shell elements (Model A) and the other uses compressive strut

29、s (Model B) as suggested in FEMA356 [10] guidelines.</p><p>　　4.1. Model A (infills modeled by shell elements)</p><p>　　Infill walls are modeled with shell elements. However, the current version

30、 of the SAP2000 computer program includes only linear shell elements and cannot account for cracking. The tensile strength of the infill walls is set equal to 26 psi, with a modulus of elasticity of 644 ksi [10]. Because

31、 the formation ofcracks has a significant effect on the stiffness of the infill walls, the following iterative procedure is used to account for crack formation:</p><p>　　(1) Assuming the infill walls are lin

32、ear and uncracked, a nonlinear time history analysis is run. Note that plastic hinges exist in the beam elements and the segments of the beam elements where moment demand exceeds the cracking moment have a reduced moment

33、 of inertia.</p><p>　　(2) The cracking pattern in the infill wall is determined by comparing stresses in the shells developed during the analysis with the tensile strength of infills.</p><p>　　(

34、3) Nodes are separated at the locations where tensile stress exceeds tensile strength. These steps are continued until the crack regions are properly modeled.</p><p>　　4.2. Model B (infills modeled by struts

35、)</p><p>　　Infill walls are replaced with compressive struts as described in FEMA 356 [10] guidelines. Orientations of the struts are determined from the deformed shape of the structure after column removal

36、and the location of openings.</p><p>　　4.3. Column removal</p><p>　　Removal of the columns is simulated with the following procedure.</p><p>　　(1) The structure is analyzed under th

37、e permanent loads and the internal forces are determined at the ends of the columns, which will be removed.</p><p>　　(2) The model is modified by removing columns A2 and A3 on the first floor. Again the stru

38、cture is statically analyzed under permanent loads. In this case, the internal forces at the ends of removed columns found in the first step are applied externally to the structure along with permanent loads. Note that t

39、he results of this analysis are identical to those of step 1.</p><p>　　(3) The equal and opposite column end forces that were applied in the second step are dynamically imposed on the ends of the removed col

40、umn within one millisecond [11] to simulate the removal of the columns, and dynamic analysis is conducted.</p><p>　　4.4. Comparison of analytical and experimental results</p><p>　　The maximum ca

41、lculated vertical displacement of the building occurs at joint A3 in the second floor. Fig. 7 shows the experimental and analytical (Model A) vertical displacements of this joint (the AEM results will be discussed in the

42、 next section). Experimental data is obtained using the recordings of three potentiometers attached to joint A3 on one of their ends, and to the ground on the other ends. The peak displacements obtained experimentally an

43、d analytically (Model A) are 0.242 in (6.1 mm)</p><p>　　Fig. 8 compares vertical displacement histories of joint A3 in the second floor estimated analytically based on Models A and B. As can be seen, modelin

44、g infills with struts (Model B) results in a maximum vertical displacement of joint A3 equal to about 0.45 in (11.4 mm), which is approximately 80% larger than the value obtained from Model A. Note that the results obtai

45、ned from Model A are in close agreement with experimental results (see Fig. 7), while Model B significantly overestimates the def</p><p>　　Fig. 9 compares the experimental and analytical (Model A) displaceme

46、nt of joint A2 in the second floor. Again, while the first peak vertical displacement obtained experimentally and analytically are in good agreement, the analytical permanent displacement under estimates the experimental

47、 value. </p><p>　　Analytically estimated deformed shapes of the structure at the maximum vertical displacement based on Model A are shown in Fig. 10 with a magnification factor of 200. The experimentally mea

48、sured deformed shape over the end regions of beams A1–A2 and A3–B3 in the second floor</p><p>　　are represented in the figure by solid lines. A total of 14 potentiometers were located at the top and bottom o

49、f the end regions of the second floor beams A1–A2 and A3–B3, which were the most critical elements in load redistribution. The beam top and corresponding bottom potentiometer recordings were used to calculate rotation be

50、tween the sections where the potentiometer ends were connected. This was done by first finding the difference between the recorded deformations at the top and bottom of </p><p>　　Analytical results of Model

51、A show that only two plastic hinges are formed indicating rebar yielding. Also, four sections that did not have negative (top) reinforcement, reached cracking moment capacities and therefore cracked. Fig. 10 shows the lo

52、cations of all the formed plastic hinges and cracks.</p><p><b>　　2，譯文： </b></p><p>　　鋼筋混凝土填充框架結構對拆除兩個相鄰的柱的響應</p><p>　　作者：Mehrdad Sasani 美國波士頓東北大學，斯奈爾400設計中心 MA02115&l

53、t;/p><p>　　收稿日期：2007年7月27日，修整后收稿日期2007年12月26日，錄用日期2008年1月24日，網(wǎng)上上傳日期2008年3月19日。</p><p><b>　　摘要：</b></p><p>　　本文是評價圣地亞哥旅館對同時拆除兩根相鄰的外柱的響應問題，圣地亞哥旅館是個6層鋼筋混凝土填充框架結構。結構的分析模型應用了有限元

54、法和以此為基礎的分析模型來計算結構的整體和局部變形。分析結果跟實驗結果非常吻合。當測量的豎向位移增加到為四分之一英寸（即6.4mm）的時候，結構就發(fā)生連續(xù)倒塌。通過實驗分析方法評價和討論隨著柱的移除而產(chǎn)生的變形沿著結構高度上的發(fā)展和荷載動態(tài)重分配。討論了軸向和彎曲的變形傳播的不同。結構橫向和縱向的三維桁架在填充墻的參與下被認為是荷載重分配的主要構件。討論了兩種潛在的脆性破壞模型（沒有拉力加強的梁的脆斷和有加筋肋的梁的擠出）。分析評價了結

55、構對額外的重力和無填充墻時的響應。</p><p>　　Elsevier有限責任公司對此文保留所有權利。</p><p><b>　　關鍵詞：</b></p><p>　　連續(xù)倒塌，荷載重分配，對荷載抵抗能力，動態(tài)響應，非線性分析，脆性破壞。</p><p><b>　　介紹：</b></p&

56、gt;<p>　　作為減小由于結構的局部損壞而造成大量傷亡的可能性措施的一部分，美國總務管理局【1】和國防部【2】出臺了一系列制度來評價結構對連續(xù)倒塌的抵抗力?！?】定義連續(xù)倒塌為，由原始單元的局部破壞在單元間的擴展最終造成結構的整體或不成比例的大部破壞。</p><p>　　通過Ellingwood 和Leyendecker【4】建議的方法，ASCE/SEI 7定義了兩種一般模型來減小結構設計時連

57、續(xù)倒塌效應產(chǎn)生的損害，它們分為直接和間接的設計方法。一般建筑規(guī)范和標準用增加結構的整體性的間接設計方法。間接設計法也應用于美國國防部的降低連續(xù)倒塌設計和未歸檔設備標準中。盡管間接設計法可以降低連續(xù)破壞的風險【6，7】，對基于此法設計的結構破壞后的表現(xiàn)的判斷是不容易實現(xiàn)的。</p><p>　　有一種基于直接設計的方法通過研究瞬間消除受載構件，比如柱子，對結構的影響來評價結構的連續(xù)倒塌。美國防部和國家事務管理局的規(guī)

58、章是要求去除一個受荷構件，考慮其影響。這樣的規(guī)范目的是評價結構的整體性和結構的一個單元出現(xiàn)嚴重的毀壞時的分荷能力。這種方法是研究結構受連續(xù)倒塌的影響的程度，但是事實上初始結構損傷的影響不止局限于某一根柱子。</p><p>　　在本論文中，應用通過實驗證實的分析結果，評價圣地亞哥旅館抵抗連續(xù)破壞的能力，實驗中瞬間移除兩個相鄰的柱子，其中一個柱是拐角柱。為了爆除這兩個柱子，將炸藥放在預先在柱子上鉆的孔里面。柱子然后

59、再用幾層保護材料包裹好，以避免爆炸時的沖擊波和碎片影響結構的其他部分。</p><p><b>　　建筑的特性</b></p><p>　　圣地亞哥旅館建造于1914年，在1924年又向南擴展了一部分，此部分包括兩個分離的結構。圖.1是從南邊看旅館的樣子。注意這張照片，旅館的第一和第三層被用黑色的布蒙了起來。這個六層的旅館是無延性的鋼筋混凝土框架結構，其中還有由空心磚

60、構成的填充外墻。擴展部分的填充墻有兩層共203mm厚。第一層的樓高為6.0m，其他樓蓋高為3.2m，頂樓高度為5.13m。圖.2為其中一個擴展部分的第二層。圖.3為對本建筑的實施計劃，即瞬間爆除一層相鄰的柱A2和A3，以評價其影響。</p><p>　　左圖：圖.1 圣地亞哥旅館的南端視角，本論文研究其中心結構 </p><p>　　右圖：圖.2 擴展結構的第二層（南端視角）</p&

61、gt;<p>　　下圖：圖.3 擬對旅館南擴展部分實施的柱的移除計劃，第一層要被移除的柱用叉號標出</p><p>　　如圖.3所示樓蓋系統(tǒng)縱向（南北向）有一個托梁。根據(jù)兩個混凝土構件受壓的實驗結果，對一個標準的混凝土柱，受壓承載力為31MPa?；炷恋膹椥阅Ａ看蟾艦?6300MPa左右。同樣，通過橫截面12.7mm的鋼筋受拉實驗，其屈服和極限抗拉強度分別為427和600MPa。鋼筋的極限變形為0.

62、17。鋼筋的彈性模量近似為200000MPa。</p><p>　　這個建筑按計劃將被爆破摧毀。作為摧毀的一個步驟，第一層和第三層的填充墻被移除。移除時上面 沒有活荷載。所有的非結構部件包括隔墻、管道設備、家具都被事先搬走了，只有梁、柱、樓板梁和在邊梁上的填充墻被留下。</p><p><b>　　傳感器布置</b></p><p>　　混凝土

63、和鋼筋的應變傳感器是用來測量梁和柱的應變變化的。線性電位計用來測量整體和局部變形?；炷翍儨y量儀常900mm，最大應變?yōu)?#177;0.02.鋼筋應變測量儀應變極限為±0.2。應變測量儀可以帶到幾百千赫茲。電位計用來測量建筑中梁沿一端的轉(zhuǎn)動和整體位移，這些以后將講到。電位計的分辨率為0.01mm，最大速度為1.0m/s，實驗中最大記錄速度為0.35m/s。</p><p><b>　　有限元

64、模型</b></p><p>　　通過有限單元法，在軟件SAP2000【8】中生成一個建筑模型。梁和柱都被抽象成Bernoulli單元。T和L型梁的翼緣計算寬度為四倍的較厚板的厚度【5】。塑性鉸可以發(fā)生在任何鋼筋可能發(fā)生屈服的地方，包括單元的端點、加筋肋分離點和彎矩的屈服點。在分析中，塑性鉸的范圍是構件高度的一半?，F(xiàn)行版本的SAP2000不能計算出單元斜裂縫的構成。為了得出正確的構件撓曲剛度，反復做以

65、下步驟：首先假設建筑的所有單元都是沒有裂縫的；然后，需要彎矩同構件的出現(xiàn)裂縫的彎矩相比較。分別降低板厚和梁的慣性矩35%，使需求彎矩大于裂縫出現(xiàn)彎矩。梁外部出現(xiàn)裂縫的正負彎矩分別為58.2knM和37.9knM。需要注意的是柱子沒有裂縫出現(xiàn)。再后，再按以上方法重新分析建筑和彎矩簡圖。重復這些步驟直到所有的裂縫區(qū)域被鑒定和用模型表示出來。除了兩端區(qū)域建筑結構里的梁上部不配筋（圖.4）。例如，梁A1-A2在距A1點305mm以后，其上部不配

66、筋（如圖.4和5）。為了確定出可能喪失撓曲強度的截面位置，將裂紋鉸布置在上部沒有配筋的可能的彎曲破壞點上。塑性鉸的撓曲強度設為于Mcr相等，當所受的彎矩達到Mcr時，該截面即發(fā)生破壞。</p><p>　　圖.4 二層的梁A3-B3和梁A1-A2詳細配筋情況</p><p>　　樓蓋系統(tǒng)有沿縱向（南北向）的次梁。圖.6所示為一典型的樓蓋的橫截面。為了計算出次梁和板的可能的非線性響應，用梁單

67、元為樓蓋建立模型。次梁按T型梁計算，翼緣的計算寬度為各自板厚的四倍【5】。選取軸2和軸3的縱梁和其之間的一個寬20英寸的梁間的格柵為板的計算模型。為了給出板沿橫向的計算模型，同樣用一個寬20英寸于橫梁平行的梁。在方形的板中其剪力流和梁單元的中的不一樣。所以其扭轉(zhuǎn)剛度取為整個截面剛度的一半【9】。</p><p>　　圖.5 梁的上部配筋彎曲位置（于梁A1-A2相似，在鄰近建筑靠近柱A1的地方）</p>

68、<p>　　圖.6 典型的樓蓋的次梁系統(tǒng) </p><p>　　圖.7 實驗和分析的第二層柱A3的豎向位移</p><p>　　建筑的2、4、5、6層有填充墻，并在門窗等開口位置有過梁，如前面提到的第1、三層的填充墻，在爆除前已經(jīng)拆掉。填充墻是用良好的空隙磚砌成的，空心磚的凈空是其總大小的一半。填充墻的平面效應增強了建筑的剛度和強度，并且影響建筑的對荷載反應即變形。如果

69、忽略墻的影響將得不到準確的建筑的剛度和強度。</p><p>　　在SAP2000中考慮了兩種填充墻的形式：一種是用平面框架模型（模型A），另一種是FEMA365【10】中建議的受壓桿件模型（模型B）。</p><p>　　4.1模型A是平面框架模型，但是，現(xiàn)行版本的SAP2000只能計算線性框架模型，不能計算裂縫的發(fā)展情況。填充墻的抗拉強度大概為26psi，彈性模量為644ksi【10】

70、。由于裂縫的發(fā)展對填充墻的剛度影響很大，重復以下步驟來計算裂縫的形成：</p><p>　?。?）假設填充墻是線性的而且沒有開裂，運行非線性歷史分析。由于梁中的塑性鉸的存在，梁中彎矩大于裂縫出現(xiàn)彎矩時候，對截面慣性矩有一個折減。</p><p>　?。?）判定填充墻出現(xiàn)的依據(jù)是看其應力于墻的抗拉強度大小關系。</p><p>　?。?）節(jié)點在拉應力大于抗拉強度的地方

71、分離。</p><p>　　重復上面的步驟直到裂縫區(qū)域被確定。</p><p>　　4.2.模型B（受壓桿件模型）</p><p>　　如FEMA356【10】所述用受壓桿件來代替填充墻，桿件的方向根據(jù)移除柱后的結構變形形式和開口位置確定。</p><p><b>　　4.3.柱的移除</b></p>&l

72、t;p>　　按以下步驟模擬柱的移除。</p><p>　　結構是在只受永久荷載下分析的，內(nèi)力在柱端測定，將隨著柱的移除而卸荷。</p><p>　　模型的建立是在移除第一層的柱A2、A3的情況下進行的。結構同樣是在永久荷載下進行靜態(tài)分析的。在此情況下，測得的柱端內(nèi)力被當成永久外部荷載施加在結構上。注意此分析結果跟第一步的分析是等價的。</p><p>　　第二

73、步中大小相等方向相反的柱端力，被瞬間施加在原柱的位置上，然后進行動態(tài)分析。</p><p>　　4.4.實驗和分析結果的比較</p><p>　　結構計算最大豎向位移在第二層的柱A3上，圖7所示為按模型A的實驗和分析的梁A3豎向位移的比較。實驗數(shù)據(jù)是用三個粘在A3兩端的傳感器記錄的。實驗和分析得到的最大位移分別是6.1mm和6.4mm，相差盡為4%。實驗和分析的位移產(chǎn)生所用時間分別為0.0

74、69S和0.066S。分析結果顯示永久位移為5.3mm，比實驗結果小14%，實驗結果為6.1mm。</p><p>　　圖.8.第二層的柱A3在模型A和B下分別沿時間的豎向位移</p><p>　　圖.8.比較了第二層的柱A3分別在模型A和B下分析的沿時間的豎向位移。由圖中可以看出，按受壓桿件模型（模型B）得出的最大豎向位移為11.4mm,比用模型A得出的結果高出約80%。在圖.7.可以看

75、出按模型A得出的結果與實驗結果是想接近的，B模型得出的結構變形過高。如果最大豎向位移偏大的話，填充墻開裂情況會更加嚴重，更偏向于受壓桿件形成，模型A和模型B得出結果差異將減小。</p><p>　　圖.9.比較了用模型A時第二層的柱A2的分析和實驗的位移值。同樣，第一次達到最大位移值的實驗和分析值非常接近，分析的永久位移值比實驗的位移值略微低些。圖.10.所示為根據(jù)模型A得出的最大豎向位移的結構變形放大200倍后

76、的情況。</p><p>　　圖.9.第二層的柱A2豎向位移實驗和分析結果比較</p><p>　　圖.10.按模型A，F(xiàn)EM分析的結構變形形式（第二層的實驗得出變形形式也給出）</p><p>　　通過實測得的變形形式在圖中也用實線標出了。在二層的梁A1-A2、A3-B3的上下端部應力重分配復雜的地方共用了14個電位計。梁上部和對應的下部電位計接在一起用來測量梁的